31 0 Broadway Square Building

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Wayne G. Lischka, P.E.

Architectural Engineer Prairie Village, Kansas

Describes

the

architectural requirements, structural con- siderations, production and erection of a totally precast prestressed concrete 17- story building in Kansas City, Missouri. A variety of precast components

including

col- umns, double tees, stairs, slabs, architectural panels and different types of beams were used. The resulting structure has met all expecta- tions regarding aesthetics, strength, durability, function and economy. The design concepts, with some modifi- cation, can be applied to other multistory buildings.

50

31 0 Broadway Square Building

A

total precast concrete ap- proach to high rise office building construction can be seen in the 310 Broadway Square Building. The 17-story building rises 12 stories above grade to a height of 190 ft (58 m) and extends five stories below grade to a depth of 63 ft ( 19 m) for a total height of 253 ft (77 m). A two- story precast concrete clad penthouse sits on top of the roof.

The structure utilizes totally precast concrete components which include all walls, both sublevel and upper level, 20, 24 and 32 in. (508, 610 and 813 mm) double tees, all beams, stairs, slabs, architectural panels and columns that range in size up to 36 x 42 in. (914 x 1067 mm) oval sections and 35 x 36 in.

(889 x 914 mm) rectangular sections.

The architects desired to create a design that visually coincides with a historic limestone building on the adjacent property while still creating a contemporary statement.

Curvilinear shapes would be required on the facade and used throughout the interior. An architectural precast concrete cladding was chosen because it could match the appearance of limestone while economically creating curves.

The owner compared a steel frame structure, a cast-in-place structure and a totally precast concrete structure. The precast concrete structure was not only ini- tially more economical to con- struct, but also, it could reduce the overall construction time, thus providing further cost savings.

As an additional cost and time savings procedure, the precast concrete structural design would be

performed while the building was under construction. This stipula- tion further complicated the project because, in the Kansas City area, it is customary that the Architect and Engineer of Record specify design requirements and then check the shop drawings to make sure these requirements are met. The precaster does the actual structU'ral design and detailing. On this project the precaster, Quinn Concrete Company, chose the structural en- gineering firm of Wayne G.

Lischka, P.E., to design and detail the precast concrete portion of the building.

Design Requirements The main architectural requirements affecting the precast concrete would be that the facade match the limestone building on the adjacent property, that the front facade be curvilinear while can- tilevering from 5 ft 6 in. to 15ft 6 in.

(1.68 to 4. 72 m), and that the columns be either circular or oval.

The main structural requirements would be that the design meet the Uniform Building Code, 1982 Edition, and that the ultimate load factors given in the ACI 318-83 Building Code Eq. 9-1, U = 1.40 + l.7L, be increased to U

=

1.80

+

1.8L, except for the vertical load carrying members.

Plans and Details

The five sublevel floors cover approximately 136,000 sq ft (12650 m²⁾ and are used for parking. A typical framing plan can be seen in Fig. 2. The floor-to-floor height is 10 ft (3.05 m). The predominant floor members are 24 in. (610 mm)

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Fig. 2. Typical sublevel framing plan.

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double tees. The increase in ultimate load factors resulted in an increase of approximately 21 percent in the loading. This posed a structural problem of too much camber in the tees and too high a release strength when the ultimate moment was met.

To reduce these problems, the~

in. (13 mm) diameter low relaxation strands were stressed to 0.617 fvu, or 25.5 kips (113 kN), rather than the standard 0.75 fvu, or 31.0 kips (138 kN). When the topping thickened or an increased loading area was required, two design changes were used. Either mild steel reinforcement was added to increase the ultimate strength, or a 32 in.

(813 mm) bridge tee was used and blocked up to the 24 in. (610 mm) depth to increase the section prop- erties. L-beams were used down the center of the building to support the sloping ramps, and inverted tee beams were used at the crossovers.

The exterior sublevel columns were 35 x 36 in. (889 x 914 mm) with a I in. (25.4 mm) clearance space used between the exterior precast concrete walls and the columns. These columns supported the upper floors and did not pick up any sublevel loading. The tees bear on haunches on the exterior walls.

The interior main columns were 36 x 42 in. (914 x 1067 mm) ovals with the ramp columns being 24 in. (610 mm) square. Columns were typi- cally cast three stories in height except for the ramp columns and the columns under the core, which were cast full height.

The reinforcement ratio in the

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columns was required to be kept below 5 percent in most cases by the Engineer of Record, with the concrete strength being increased from 5000 to 6000 psi (34.5 to 41.4 MPa) when this limit was exceeded. Although the reinforcement ratios were not increased on this project, it should be noted that research seems to indicate that when lap splices are not used in the reinforcing bars and when a top and bottom plate is used, it can be jus- tified to use a reinforcement ratio as high as 17 percent. A review of the current research should be made prior to an engineer exceed- ing the 8 percent limit.

The exterior sublevel walls on this project were all precast concrete, which resulted in a major cost and time savings over a cast- in-place wall. Fig. 3 shows a partial

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elevation of how the sublevel walls were used. The walls span vertically from floor to floor so the vertical joint locations are not structurally critical. The walls were cast 12 in. (305 mm) thick with the standard width being 10ft (3.05 m). The walls were designed as column strips with unsymmetrical reinforcement.

Fig. 4 shows the typical horizontal and vertical joints. The vertical joints were placed to the side of the columns rather than at the center of the grids so that if a leak does occur, it can be seen and more easily repaired. The horizontal joints were placed above the top of the tees so that the horizontal loads could be transferred into the diaphragms through direct bearing rather than requiring a mechanical connection between panels.

HORIZONTAL JOINT DETAIL VERTICAL JOINT DETAIL

Fig. 4. Typical sublevel joints.

52 PC! JOURNAL

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It should be noted that a mechanical connection was designed to transfer the loading and is shown in the detail, but idealistically it is only there for redundancy. The sublevel walls were also used as retaining walls at grade level in some locations. In these cases the walls were broken into vertical elements, as shown in Fig. 4. This resulted in a considerable footing cost savings.

The upper levels contain approximately 290,000 sq ft (26970 m²) of office space. An 18 ft (5.49 m) floor-to-floor height was used on the ground floor and a 13ft (3.96 m) height was used on the standard floors. A typical floor can be seen in Fig. 1. The standard floor tees used on the upper levels were 20 in.

(508 mm) deep double tees.

The predominant feature on these floors is the curvilinear facade on the east side. The tees cantilever out 15ft 6 in. ( 4. 72 m) at the ends to 5 ft 6 in. ( 1.68 m) toward the center while creating a double reverse curve. To create the cantilevers, 24 in. (610 mm) double tees were used. In the outside bays a 32 in. (813 mm) bridge section was blocked up to 24 in. (610 mm) to increase the stiffness. A typical cantilever tee can be seen in Fig. 5.

Mild steel reinforcement was placed in the bottom of the stems at the cantilever end to resist stripping, handling and erection loads.

Mild steel was also placed in the top at these ends to control differ- ential camber. On the outside bays the strand holdup point was varied to control camber as the cantilever spans changed from 15ft 6 in. to 10 ft 6 in. (4.72 to 3.20 m). On the in- side bays no holdup point was used, and the strand holddown point was varied to control the cantilever camber.

A soffit beam was used to provide support for the cantilever sections as seen in Fig. 6. The lower portion of the beam was prestressed concentrically in the plant with stirrups extending out that projected through the flanges of the tees and into the topping. Mild steel was extended through voids in the stems of the tees and anchored into

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BEARING PAD

GROUT SOLID (BY ERECTOR)

Fig. 7. Horizontal core joint.

the columns to provide shear friction steel. Steel was also placed in the topping under the stirrups, and the remaining portion of the beam was poured in the field.

Twin cores were designed to resist all the lateral loads as tubes.

The joint design was critical to create a monolithic core. Horizon- tal joints were designed using the NMB splice sleeve system. A typical joint can be seen in Fig. 7. The vertical joints were designed as grooved joints with a typical joint being shown in Fig. 8. The joint strengths were designed using the shear friction method. Steel plates and studs were used when the joint was required on the form face of the panel. Tolerances are critical on this type of design since the panels must interlace in the vertical joint and then slide down to com-

plete the horizontal joint.

Diaphragm Design

The sublevel diaphragm required an unusual method of design since the floors were parking ramps. The maximum ultimate loading on the diaphragm was 21.6 kips per lineal

54

CROUT BLOCKOUT

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BY ERECTOR

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Fig. 8. Vertical core joint.

ft (315 kN/m). In the north-south direction the loading was assumed to be transferred up and down the ramps in compression. Forces were easily resisted in this direction, but in the east-west direction the sloping ramps resulted in a beam with

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end supports 179 ft (55 m) apart.

The original design concept short- ened the lateral beam design by using the vertical core walls at Grids 3, 4 and 4.8 to transfer the load between ramps. A keyed joint system with Lenton dowels at 24

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Fig. 9. Shear block detail.

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in. (610 mm) on center in all vertical walls in the sublevel was specified. This proved to be too costly.

In their place, a series of shear blocks was used (Fig. 9). In the shear block concept, the shear was transferred through both the double tees and the topping. Mechanical connections between the tees were designed using standard design concepts and spacings. The remainder of the shear had to be transferred through the topping.

At the transfer walls a shear

block was formed between the tee stems. This block was designed to transfer the shear out of both the tees and the topping and transfer it into the walls through bearing. The shear friction concept was used by casting studded plates into the side of the stems and after stripping, deformed anchors were welded to the plates. Stirrups circled this shear friction steel and the main reinforcing steel from the topping.

The block was poured with the topping.

Bending stresses and the chord

Fig. 10. Computer output showing stress contours in diaphragm.

Fig. 11. Computer output showing bending stress contours in large core.

design proved not to be a problem, but shear exceeded the standards set in the ACI Code. The concept that arching could exist in the topping and hence, shear would not be a problem, was proposed and considered. Unfortunately, published research on this method did not appear to be readily available and the fixity required by the supports of an arch seemed questionable in a diaphragm.

A finite element analysis was run for the southeast corner diaphragm section. One can see on the stress output (Fig. 1 0) that arching does not appear to take place, but the diaphragm follows the design concepts used in deep beam theory.

The shear stresses indicated that the majority of the diaphragm was within the limits of the ACI Code and that only two localized areas exceeded acceptable limits. These areas were thickened and special reinforcement added.

Lateral Analysis

The building was designed to resist lateral loads by two cores. The centroid of the cores did not coin- cide with the centroid of the building, and the stiffness changed greatly as walls changed to columns in the sublevels. Only the small core would have two small walls continue down through the sublevels in the north-south direction. This flexibility of the base and changing stiffness made a standard two-dimensional analysis suspect.

It was decided to design the cores as tubes using a three-dimensional finite element analysis.

The analysis was done in house on an IBM PS/2 Model 80 computer using a program called SCADA (Structural Computer Aided Design & Analysis). The four node plane stress element was used to model the walls and the 3D beam element was used on the columns. A course grid was used due to the complexity of the problem.

About 800 nodes existed. A run with both multipoint constraints and a lateral beam was used to model the diaphragms to tie the two cores together.

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Fig. 12. Computer output showing bending stress contours in small core.

The lateral analysis model was 17 stories tall with five stories extending below grade. The structure was modeled assuming that restraint would be provided by the diaphragms and backfill in the sublevels. Lateral restraints were provided in Sublevels 1-5. Ten load cases were run and covered ACI Eq. 9-2, U = 0.75 (1.40 + 1.7L + 1.7W) and ACI Eq. 9-3, U = 0.90 + 1.3W. The built-in graphics report- er was used to interpret the output file, which at 4.6 Mbytes would have been practically unusable without such a secondary program.

The bending stresses from the lateral load being applied in the north-south direction under Eq. 9.3 can be seen in Fig. II. Window I sits on top of Window 2. The cou- pled shear walls of the main core rest on four columns as does the right corner of the small core. The columns do not show up in the stress output since they are 30 beam elements, but they are in- cluded in the model.

It is interesting to note that ten- sion not only exists at the top of the main core, but exists in the small core when the wall sidesteps between upper Level 2 and Sublevel I. Also note that the maximum bending stresses occur in the upper sublevels where lateral restraint is first provided. The bending stresses from the lateral load being

56

applied in the east-west direction under Eq. 9.2 were similar. The maximum bending stresses again occur in the upper sublevels.

The shear stresses were most critical with the lateral load applied in the north-south direction. Only two small walls on the small core carried all the way down to the base of the structure. Stresses in one wall can be seen in Fig. 12. The shears were transferred into the diaphragms at Sublevel 1 as ex- pected, and the stresses were rela-

Fig. 14. View showing section of oval form.

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Fig. 15. Typical Z-beam and soffit panel.

tively low. Structurally, what is interesting here is that the lower levels had to be restrained to keep the stresses low. A structural concept of flexible bases and shifting shear walls worked in this structure because it did extend below grade.

Production

To produce a large scale precast concrete project requires that the precaster be prepared to handle some unusual forming requirements. Many unusual forms were used on this project, but two sig- nificant types of special forms existed - the round or oval columns and the Z-beams. The majority of the components on this project could be cast on standard beds or with standard architectural forming techniques.

The average column on this project is three stories tall, or 39ft (12 m) in length. Pouring the columns vertically was not practical, and trying to form a rounded surface by hand finishing was considered impossible. The solution was to take a circular form and separate it vertically into two halves separated by a 6 in. (152 mm) flat section.

Typical form layouts can be seen in Fig. 13. This created the appear-

ance of an oval column, although it was not a true oval.

To simplify connections at the floors, the columns were cast rectangular between ceiling and floor.

This allowed corbels to be changed in flat plate form sides rather than in curved areas. This reduced the cost of form changes. One section of an oval form can be seen in Fig.

14. The concrete was poured through the 6 in. (152 mm) opening at the top and the sides slid away for stripping.

Other components of the project that required special forming techniques were the Z-beams. The use of a typical Z-beam can be seen in Fig. 15. These beams were used to create cantilevers or circular formed openings on the interior floors for aesthetic effects. These

ERECTION SEQUENCE - SUBLEVELS

ERECTION SEQUENCE - UPPER LEVELS

1-SUBLEVEL 5 TO 4

2-SUBLEVEL 5 TO GROUND INCLUDING CORE TO GROUND 3- CORE: GROUND TO LEVEL 9

4-SUBLEVEL 5 TO GROUND REMOVE CRANE FROM EXCAVATION

5-SUBLEVEL 5 TO GROUND & AREA #1 TO GROUND 6-CORE: LEVEL 9 TO ROOF

7- SET TEES AND SHORING, POUR SOFFIT BEAM, ALTERNATE FROM A TO B 8-SIMULTANEOUS W/ #7 USING SECOND CRANE

9- PENTHOUSE STRUCTURE (NOT SHOWN)

Fig. 16. Erection sequence in sublevels (top) and upper levels (bottom).

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beams were all precast with mild steel reinforcement rather than prestressed steel because of the formwork required.

Erection

A team effort between the precaster, erector and engineer was required from the conceptual design of the structure through the final detailing. Erection would begin about 63 ft (19 m) below grade and would end with the precast concrete penthouse panels being erected about 190 ft (58 m) above grade. A totally precast concrete project, 253 ft (77 m) tall with individual pieces weighing as much as 72,000 lb (320 kN) in the lower core and as much as 53,000 lb (236 kN) on the roof parapet, required erection to be a major considera- tion.

Erection was further complicated by the building being located on a major downtown intersection.

The city would not allow any street to be closed and would only allow one lane to be used for erection and delivery between the hours of 9:00 a.m. and 4:00 p.m. until the end of the project. Due to the weight of the pieces and the maneuverability required by the site, the erector chose to use a Manitowoc 4100 WV S-2 230 ton (209 t) crane with a 240 ft (73 m) main boom and a 60ft (18 m) jib over a tower crane.

Erection can be viewed in nine distinct phases (Fig. 16). The building extends five sublevels below grade and 12 stories above grade, plus it has a 27ft (8.23 m) tall precast concrete clad penthouse.

Excavation was completed at the site with the exception of a ramp into the excavation. The crane was walked down this ramp, and the excavation equipment removed the ramp as they exited from the hole.

Figs. I 7 through 23 show various construction stages of the building. Erection began in the southeast corner of the building with the first bay being erected up only two sublevel stories. Erection then proceeded in Phase 2 down the east half of the structure in a counter- clockwise rotation with the

58

Fig. 17. Start of erection showing partially completed columns.

Fig. 18. Crane in position to erect superstructure.

structure being erected from Sub- level 5 to the ground floor. The core was erected to the ground level in Phase 2. After proceeding to Grid 2 on the west half of the building, erection to grade level ceased. Phase 3 then began with the core being erected to Level 9 and the core columns extending to Level 10. Level 9 is 14 stories above the crane position at this point.

The core length was chosen to extend between Grids 2 and 5 in order to erect both precast con-

crete stair towers. This allowed easy access for the construction crews. Erection of the core also allowed work by other trades to proceed, resulting in an earlier project completion date.

Once the tower was erected as high as possible from the base of the excavation, Phase 4 began with the erection of the remainder of the sublevel structure up to ground level minus the last two to three bays. It was important to note that the tees between Grids 3 and 4 were detailed as double tees on the

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Fig. 19. Shoring of slab and double tees.

Fig. 20. Lower stories partially completed.

east side and single tees on the west side.

The core pieces were detailed as large as possible to reduce the overall number of pieces. This in turn required that the crane be po- sitioned close to reduce the lift angle. Therefore, the west side tees were detailed to be brought into the excavation at an angle and slid under the core between the columns. Only 2 ft (0.61 m) of clearance space existed and the tees extended approximately 15 ft (4.57 m) under the core. The lifting han-

dies on the center side of these tees were located about 15 ft (4.57 m) from the end, and mild reinforcing steel was added to resist the erection loading.

Phase 4 ended on a Friday with the 230 ton (209 t) crane being located in the southwest corner of the excavation. At this point the crane would be dis"assembled, lifted from the excavation and reassembled over the weekend so erection could proceed Monday morning. The street to the south was blocked for the weekend. The boom was then

Fig. 21. Construction progress.

Fig. 22. Double tees in place.

lowered over the Phase I area which was not erected to ground level for this purpose.

The entire boom was removed in one piece with the cables still in place to reduce assembly time.

Two additional cranes were used for this purpose. The counter- weights were then removed, fol- lowed by the cab and then the treads and base assembly. Disas- sembly was completed on Satur- day, and reassembly was completed on Sunday.

Erection on Phase 5 then com-

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..(,

Fig. 23. Top view of building construction.

Fig. 24. Front view of building.

menced and completed the sublevel structure. The core was then completed on the 13th level (roof) above grade. Erection on Phases 7 and 8 then proceeded simultaneously with two cranes. Erection on Phase 7 was critical since it in- cluded the soffit beams which had to be poured in the field. Each half (7 A and 7B) took 2!1 days to erect.

The soffit beams were poured and heated from below.

Shoring was removed at about the same schedule, which required the beams to be at design strength

60

quickly. The architectural panels on the west side were placed on Phase 7 after the structure was completed. The architectural panels on Phase 8 were erected simultaneously. The penthouse panels were erected in Phase 9 after the mechanical equipment was placed in the penthouse.

When erection was complete, 2936 pieces had been erected with an average of 14 pieces per crane per day. Seventeen stories, five below and 12 above grade, and an additional two-story penthouse

Fig. 25. Side view of building.

Fig. 26. Complete building.

were erected in approximately 210 working crane days. Actually, erection took less than 210 days due to two cranes being used for a portion of the construction.

Figs. 24 to 26 show different views of the completed building.

The building was essentially completed in October 1989. At the time of opening, the facility had a 75 percent occupancy.

The total cost of the project was

$25 million with the precast concrete portion of the work (architectural plus structural plus erec-

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tion) amounting to $5,920,000. Based on a working area of 437,800 sq ft (40715 m²^),the precast concrete unit cost was $13.52 per sq ft.

Concluding Remarks

The 310 Broadway Square Building showed that precast concrete is a practical solution in high rise office construction. It is be- lieved that the basic design concepts used on this project could be used on a building project in any location. Some of the connections used here would need to be changed in higher seismic zones, but with the technology that is now available to the designer, this should no longer be considered a limitation.

Credits

Owner: Broadway Square Partners, Inc., Kansas City, Missouri. Architect: PBNI Architects, Inc.,

Kansas City, Missouri.

Engineer of Record: Bob D.

Campbell & Company, Kansas City, Missouri.

Precast Structural Engineer: Wayne G. Lischka, P.E., Prairie Village, Kansas.

Precast Concrete Manufacturer:

Quinn Concrete Company, Mar- shall, Missouri.

General Contractor: Winn-Senter Construction Co., Kansas City, Missouri.

Building Erector: Building Erec- tion Services Inc., Olathe, Kan- sas.

REFERENCES

I. ACI Committee 318, "Building Code Requirements for Reinforced Concrete (ACI 318-83)," and

"Commentary on Building Code

Requirements (ACI 318-83),'' Amer- ican Concrete Institute, Detroit, MI, 1983.

2. Klein, Gary J., "Design of Spandrel Beams," PCI JOURNAL, V. 31, No.5, September-October 1986, pp. 76-125.

3. Mattock, Alan H., and Theryo, Teddy S., "Strength of Precast Pre- stressed Concrete Members with Dapped Ends," PCI JOURNAL, V.

31, No.5, September-October 1986,

pp. 58-75.

4. PCI Committee on Tolerances,

"Tolerances for Precast and Pre- stressed Concrete," PCI JOUR- NAL, V. 30, No.1, January-Febru- ary 1985, pp. 26-112.

5. PC/ Design Handbook -Precast and Prestressed Concrete, Third Edition, Prestressed Concrete In- stitute, Chicago, IL, 1985.

6. SCADA Users Manuals, Scada Systems Corporation, Los Angeles, CA, 1985.